canadian earthquake design

huangyhg

canadian earthquake design
using the national building code of canada, we calculate a value of ve, which represents the force that a building would experience if it were designed to respond elastically.  the code then allows us to reduce this load by a factor of r, which is a number that represents the ability of the structural brace to respond in a ductile manner.  
the idea behind all of this is to have the ductile brace yield at a lower force than that which can be delivered by an earthquake.  by doing so, the remaining components of the structure, such as the roof diaphragm and the footings, etc. only have to be designed for a little bit more than the force that it takes to yield the brace.  
however, if the brace is designed too strong, the roof diaphragm and the footings must also be increased accordingly.
the problem that i have, is that i can't figure out how to get an ebf, a moment resisting frame, or a cbf to yield at exactly ve*.6/r for a single storey building.  there always seems to be some other criteria that makes the frame stronger, such as deflection limits, etc.  
has anybody else had this problem?
dave tipler
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welcome to what they call "capacity design".
you might want to try several bays of tension-only flat bar bracing, which gives you ability to just size it strong enough, if it's allowed to be considered a ductile concentrically braced frame.
i haven't done any building work for a couple of years, but i have had the same problem.
tip: in approx 8 years of designing buildings in canada, we never used an r=4 moment frame or an r=3 braced frame.  we always went for nominal ductility.
thanks trainguy
your response is consistent with the recommendations of the cisc.  however, as we know, tension only plate braces have been found to be terrible bracing   
if you are talking about one story buildings, using steel braced frames, i presume that the structure is also steel.  if this is so, would wind not govern your design??  therefore you don't have the concern for the plates to "stretch" after the first cycle. (this is also the first that i've heard of a problem with plate braces, if you have more info/reports i'd be interested in reading them).
one issue i had with using flat bar bracing was with a drawings examiner (from an unnamed canadian city) that didn't want flat bar used because it didn't satisfy clause 10.2.2 from s16.1 (m89).  in a flat bar the slenderness ratio is well over 300, and he wouldn't let it slide past even though it's a recommended clause.  to save time we changed all the flat bars to angles.  have you ever had a similar problem??
i'm almost certain that the only reason kl/r has to be kept under 300 is to prevent your brace from vibrating excessively. you could always do something sneaky like provide attachment to a specific girt at midspan, or more spans, making sure your detail allows for axial displacement, although i've never actually designed that specific detail.
i've never heard of plate braces being so bad, i thought that you want them to yield...
no, wind does not govern in a lot of cases.  if you are talking about british columnbia, for sure, earthquake will govern.  if you are talking about quebec and ottawa, most likely, earthquake will govern.  if you are talking about large 'big box' buildings, earthquake will probably govern even in toronto.
trainguy
you slipped in your response to cottageguy while i was typing mine.  you are correct in saying that you do want the braces to yield to limit the force.  however, for tension only braces, after they yield during the first cycle, (and yield further after successive cycles), there is an associated impact loading on the braces during load reversals because there is nothing restraining the building until the building deflects to the point where the tension brace tightens up again.  if you look at the hysteresis loops for a tension only brace, you will see that they are very poor.  as a matter of fact, tension only braces are prohibited in many parts of north america where high seismic activity is expected, according to michel bruneau, the author of the book that i have previously mentioned.
this is also why the csa s16 further penalizes tension only braces by adding a load factor of 1.1 to the earthquake force.
i've been told by the experts that my understanding of this topic is excellent.  so why can't i figure out how to  utilize an r factor greater than 1 or 2?  help!
cheers!
i expect our american colleagues to be aware of the r factor for seismic design, since, the canadian s16 provisions have been somewhat based on seaoc (california) provisions circa 1990. (i was involved in the draft of seismic design clauses for s16).
my additional trouble with the requirements:
1) why do we need a u  (=0.6) factor? why can't ve be implicitly calibrated?
2) why do we not have more input from our mech. eng friends regarding the design of inexpensive dampers (such as friction(pall et al) instead of relying on this somewhat violent concept of yielding and buckling braces?
regarding the u factor, they apparently replacing this with two factors in the 2000 national building code.  we're all in for some fun!  as far as the reason why we need the factors; the reason is that the seismologists that do all the earthquake research think that we should work with the numbers that they develop from their probability studies.  i was told this in an earthquake seminar by those very people.  crazy but true!
you seem to be thinking along the same lines as myself regarding different types of systems.  i thought about a friction type connection, however, i think the reason why friction type slip connections are not being talked about is because there is about a 25% drop between static friction and kinetic friction.  it did seem to me, however, that it still might be the way to go.  i have talked to some mechanical engineers about load limiting devices, and they started talking about pressurized fluids.  ideally, what we want is something that has very little deformation up to a certain maximum load, and then displaces under a constant load.  it can't be a spring because force in a spring constantly increase with deformation.  also, deformations will probably be too much.  let's design one of these dampers.  maybe we'll get rich!
cheers!