huangyhg

tabular diaphragm values for wind design??
in the "pole barn" industry, we depend very heavily on metal-clad roof and wall diaphragms. for those not familiar with this method of construction, we use mechanically laminated dimensional lumber columns in the ground typically 8' o.c. wood trusses are typically 8' o.c. as well. the framing on the roof then is 2x4 purlins (on edge) and on the walls it is 2x4's flat. corrugated steel panels are screwed to this wood framing to provide a diaphragm assembly.
an associate and i were having a discussion about whether or not we can take the tabular shear values of tested diaphragm assemblies and increase their capacity by a 4/3 or 1.6 factor. i'm kind of torn. i read a thread about the 4/3 factor, but i believe that is more of a reduction in loads due to the likelihood of a combination of two or more events happening simultaneously. chap 16 of the ibc states clearly that "increases in allowable stresses specified in the appropriate materials section of this code or referenced standard shall not be used with the load combinations of section 1605.3.1 (ibc 2000) except that duration of load increase shall be permitted in accordance with chap 23."
the ibc does not publish steel-clad diaphragms like it does wood sheathing clad diaphragms, but does reference a standard called diaphragm design of metal-clad, post-frame rectangular buildings. nowhere in that document does it give any values or state that tabular diaphragm values from tested assemblies may or may not be increased.
for wood-on-wood diaphragms, the ibc permits a 40% increase in tabular values for wind design. does it seem reasonable that since the metal-cladding is attached to wood framing that i can use the load duration factors for wood design (nds), or 1.4 like the ibc allows for wood panel diaphragms, or even a 1.33 increase? any thoughts would be appreciated.
thank you.

i have never increased the values in the post-frame building design manual.
daveatkins
whoa!! there's somebody out there that speaks the same language!! thanks dave for your response. you'll note in that manual that it's somewhat out of date, but they mention that in certain codes (old ones) that you are permited to take an increase in the allowable shear values.
i know you said that you've never done that, and now i'm wondering why not? does today's code prohibit us from doing it? did you just want to be extra safe? do you do lots of pf buildings?
thanks.
well, looking at the load side of the equation, if you use kd = 0.85 (i always do) then you cannot use an allowable stress increase of 1.33 for masonry or steel design. however, you can use a load duration factor of 1.6 for wood design. if we knew that the values in the tables are governed by failure in the wood, then i would be comfortable with using a 1.6 increase. but if it is the steel roof deck that controls the table values, no increase would be allowed. i don't know what controls the table values, so i am not comfortable using a 1.6 increase.
daveatkins
gotta love eng-tips! i always wondered what the basis was for pole building designs (i've not designed them, just spec'd them), and i never new that manual existed! thanks.
part of the thinking for the allowable load increase, at least in the lateral forces, is the increase in allowable deflection for pole structures since most are for agricultural use, not dwellings for human habitation. more allowable deflection for the same section implies more load applied.
mike mccann
mccann engineering
let me thrown in a new twist here. the 10psf minimum wind pressures on the vertical projected plane of a buildings often governs my lateral design. frankly, it seems quite ridiculous to have one seemingly arbitrary number trump (by a large margin) a whole bunch of calculations to determine what the actual wind loads are on a building, but that is the code as i understand it. can anybody make me feel any better about that provision??
seeing then that we have to live with that 10psf loading, which doesn't account for the kd (wind directionality factor) or any .75 factor or such like, is there anything that i can do to either the load side or the material side of the equation (legitimitely) to counteract those (in my opinion) unnecessarily high loads?
just for example, on the buildings that we typically do, the 10psf load is equivalent to a calculated 110mph wind speed as opposed to the 90mph windspeed. furthermore, if we have a building that is category 1, such as a storage facility, the 10psf minimum totally ignores that and makes the comparable "effective" wind speed 115mph or so.
how do other practicing engineers apply the 10psf minimum load? i'd appreciate hearing some discussion on that.
thank you.
i try to design for the minimum 10 psf wind load, if that governs. if i find something that seems overly conservative, for example, a purlin to purlin connection that requires lag screws instead of nails, i have on occasion used the "actual" wind load just for that part of the design.
daveatkins
re
thank you mike,
i understand that the roof loads go away in the vertical direction, but the 10psf still needs to be applied horizontally to the vertical roof projection. that really is what "hurts."
i understand as well that it never hurts to have some extra safety factor built in, but don't we have enough already with diaphragm values with a 2.5x safety factor, along with all of the other published values that an engineer uses in design with their associated safety factors?
i appreciate each of the inputs to this thread. anybody else care to share your perspective on this? has everybody else just accepted this 10psf thing and i'm still fighting it years later?
thanks.
i agree that the 10 psf on the vertical projection of the roof is arbitrarily conservative. with some buildings, the "actual" wind load on the roof is in the reverse direction--opposite to the wind load on the walls! that is why i have done what i wrote in my earlier response, on occasion.
daveatkins

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